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Integrity of Structural Steel After Exposure to Fire R. H. R. TIDE INTRODUCTION welded connections will also be discussed because of their Perceptions of fire vary depending on the circumstance to impact on individual member behavior. To analyze these which the individual is exposed. Controlled fires are rarely
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  Integrity of Structural Steel AfterExposure to Fire R. H. R. TIDE INTRODUCTION P erceptions of fire vary depending on the circumstance towhich the individual is exposed. Controlled fires are rarelygiven much thought in our daily experiences. Uncontrolledfires, with the specter of buildings collapsing, the implieddamage, potential injury and loss of life have created a verynegative image. A negative connotation often exists thatanything exposed to fire and heated to a high temperaturemust be damaged, regardless of the appearance of thestructural members.Exposure to fire will subject structural steel to thermallyinduced environmental conditions that may alter its properties. Assessing these altered properties requires acombined knowledge of metallurgical and structural behavior as the fire raises the steel temperature and the steel later cools. Knowledge of steel properties and behavior developedfrom basic steel production, thermal cutting, thermal or mechanical straightening (or curving), heat-treating andwelding provides the requisite information.Fire represents a transfer of energy from a stablecondition to a transient condition as combustion occurs;examples are the burning of warehouse contents, officefurniture, books, filing cabinet contents, or other material.During this process, the steel temporarily absorbs asignificant amount of thermal energy. Subsequently, the steelstructure returns either to a stable or unstable condition after cooling to ambient temperatures. During this cycle,individual members may become badly bent or damagedwithout affecting the stability of the whole structure. It is possible to predict the range of temperatures that a particular steel member of a building experienced during a fire usingcurrent heat transfer theories. Damaged members areindicative of energy redistribution within the member itself and possibly the whole structure. Assessing overall structuralstability can proceed after the condition of the individualmembers has been established.This paper primarily addresses the structural integrity of individual members and not the structure as a whole. Theeffects of elevated temperature on high-strength bolts andR. H. R. Tide is a senior consultant with Wiss, Janney,Elstner Associates, Northbrook, Illinois, and is a member of the AISC Committee on Specifications.welded connections will also be discussed because of their impact on individual member behavior. To analyze theseeffects, it is important to review the process and temperatures by which steel is manufactured. This review is followed byan examination of previous work on steel shapes andconnections exposed to elevated temperatures. Thisknowledge is then applied to evaluate the structural behavior of individual members.Following the background discussion, means of evaluating fire-damaged steel will be presented. Various test procedures will be reviewed to examine their ability todetermine metallurgical or structural degradation of the steel properties. These tests commonly include visual observationsand measurements, surface hardness readings, residual stressmeasurements, metallographic sectioning and testing for chemical and physical properties. BACKGROUND Prior to evaluating the integrity of fire exposed structuralsteel, it is appropriate to examine some relevant backgroundinformation. A brief explanation of steel fire testing and protection (fireproofing) is given by AISC (1969). TheASTM E119 fire test referenced in this document specifiestime-temperature and loading conditions that must besatisfied. After test completion, the structural steel assemblymust satisfy certain acceptance criteria. One such acceptancecriterion for structural steel columns is: Regard the test as successful if the transmission of heat through the protection during the period of fireexposure for which classification is desired does notraise the average (arithmetical) temperature of thesteel at any one of the four levels above 1000°F(538°C), or does not raise the temperature above1200°F (649°C) at any one of the measured points. Similarly, for steel beam assemblies, a representativeacceptance criterion is: For steel beams the temperature of the steel shallnot have exceeded 1300°F (704°C) at any locationduring the classification period nor shall the averagetemperature recorded by four thermocouples at anysection have exceeded 1100°F (593°C) during this period. 26ENGINEERING JOURNAL / FIRST QUARTER / 1998  © 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without the written permission of the publisher.  Unfortunately, local or overall damage and deformation,which can occur in a fire, are not addressed in these criteria.The absence of a statement acknowledging a level of deformation, which does not impact on the structural capacityof the steel member, can imply that virtually any deformationresults in steel not acceptable for future use. Thisinterpretation oftentimes results in rejection of otherwisemetallurgically and structurally sound steel.Because of the economics involved in protection before afire or salvaging steel after a fire, research and investigation programs continue, Thomas (1992). Selected fire testliterature is referenced because relevant steel temperatures or conditions are presented. Similarly, early repair and salvagecase studies were located, which form the current knowledgeand experience database. Elevated temperature tests have been conducted on structural steel to determine behavior under prescribed fire time-temperature conditions, protectionconditions, ventilation and distance from the center of thefire, Butcher (1966, 1967a, 1967b), Jeanes (1986), Law(1967) and Thomas (1992). Depending on the test variable,the steel temperature may or may not have exceeded 1200degrees F. In this work, the visual appearance andmetallurgical or structural properties were not directlyaddressed. The primary test variable was whether or not thesteel members passed a prescribed time-temperature-deflection criterion. It is inferred, based on previous researchand production knowledge, that no significant metallurgicalchanges occurred.Steel industry firms experienced in steel production andfabrication procedures have routinely repaired or straightened fire damaged steel. Early examples referenced inthe literature include those by Corbit (1950), Dill (1960) andStitt (1964). Unfortunately very little technical informationwas reported to indicate time-temperature and metallurgicalconditions. However, Wildt (1972) formalized theseexperiences, coupled with more current research and case  Fig. 1. Representative phase diagram for structural steel. studies, without providing the necessary backgroundinformation.Based on the early fire research, the following pointsshould be recognized when evaluating steel structuresexposed to fire: ã The steel temperature is significantly reduced as thedistance from the center of the fire increases. ã Ventilation of a fire compartment greatly reduces thetemperature of steel at distances away from the center of the fire. ã The temperature of fire-protected steel in the vicinityof the fire is significantly less than unprotected steelat the same location. REVIEW OF STEEL PRODUCTIONAND PROPERTIESStructural Shapes Basic steel production temperatures and structural steel properties at elevated temperatures with respect to individualmember behavior form the basis of evaluating fire-damagedsteel. Generally, the steel ingot temperature prior to rollingranges from 1900 degrees F to 2300 degrees F depending oningot size, shape, chemistry and the size of the final product,USS (1985). For most hot rolled shapes, final rolling occurswhen the steel is approximately 1600 degrees F or higher depending on mill procedures. For members other thanstructural shapes and plates, finished steel surfaces can be produced at temperatures as low as 1400 degrees F. Re-heating steel to elevated temperatures is a well known phenomenon, Dill (1960) and USS (1985). Stress relievingtemperatures range from 1100–1200 degrees F, andannealing and normalizing temperatures are 1500–1600degrees F.The phase diagram, Neely (1979), shown in Figure 1indicates that with normal structural steel chemicalcomposition, rolling occurs when the grain structure isclassified as austenite. Under normal conditions the austeniteis transformed predominately into ferrite and pearlite as thesteel cools to ambient temperatures. Depending on thechemical composition, material thickness and rolling pressures, the resulting structural steel shape has the ASTMspecified range of yield strength (  F   y ), tensile strength (  F  u )and modulus of elasticity (  E  ). For all practical purposes, thelatter property is a constant value at ambient temperatures for structural steel regardless of chemical composition androlling processes.Once rolling of the structural plate or shape is complete,the member is placed on a cooling bed. Residual stressesdevelop because of the differential cooling rate. Dependingon subsequent cold working operations and fabrication procedures, an unknown magnitude and distribution of residual stresses occur. Residual stress development duringsteel manufacture has been addressed by Tide (1985, 1987,1989), Hall (1969), SSRC (1988) and others, and will beaddressed in a later section.ENGINEERING JOURNAL / FIRST QUARTER / 199827  © 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without the written permission of the publisher.  Heating of steel, such as during a fire, results inmetallurgical changes that are predominantly temporary,although some may be permanent. Tests indicate that as thestructural steel is heated to 1200 degrees F, the commonstructural properties  F   y ,  F  u  and  E   decrease as illustrated inFigures 2, 3 and 4, respectively. These curves representASTM A36 steel, but in a general sense can represent allcommon structural steels, USS (1981) and Lay (1982). Thecoefficient of thermal expansion is also temperaturedependent, AISC (1989): α =6.5 × 10  –6  in./in. per degree F for temperatures lessthan 100 degrees=(6.1 + 0.0019 T  ) × 10  –6  in./in. for temperatures between 100 and 1200 degrees Fwhere: T  =temperature, degrees FAbove 1200 degrees F, steel properties decrease sodramatically that they are of no structural interest.Because of their relatively low carbon and other alloyingcontent, structural steels usually regain close to 100 percentof their pre-heated properties provided the steel temperaturedoes not exceed approximately 1330 degrees F, Avent(1992), Kirby (1986), Hineman (1983) and Smith (1980).This temperature coincides with a boundary condition shownin the phase diagram of Figure 1. Any temperature rise between 1330 degrees F and the minimum temperature for hot rolled structural shapes (1600 degrees F) has minimumimpact on the structural properties once the steel has cooledto ambient temperatures. Restricting thermal heating to 1200degrees F, AISC (1989), provides a safety factor againstdegradation of the metallurgical and physical properties.Both AASHTO (1992) and AREA (1989) also adopt the  Fig. 2. Representative yield strength— temperature diagram for structural steel. 1200 degree F temperature limit for heating steel duringrepair operations, with minor variations. Elevatedtemperature creep is an important consideration for steelsupporting loads and subjected to elevated temperatures for extended time periods.Heat-treated high strength alloy steels require specialconsideration after exposure to fire conditions. These steelsare seldom used in building construction, but some bridgeshave been constructed with various proprietary products.Whenever encountered, the concepts presented in this paper must be modified accordingly. Connections High strength bolts warrant a separate mention because of their special manufacturing requirements. The increasedstrength of ASTM A325 and A490 bolts, above soft boltstrength, is obtained by combining chemical compositionwith a heat treating process. To minimize undesirable behavior, the hardness of the finished bolt is restricted to afairly narrow range. As a result, the manufacturer carefullychooses the chemistry and heat process for the grade and sizeof ASTM A325 or A490 bolt. Because of relatively similar chemistry, the phase diagram shown in Figure 1 is stillrepresentative for high strength bolts.In general, the bolt is heated to slightly above 1600degrees F for a period of time to fully austenitize the grainstructure. The bolt is then fully quenched to minimize anygrain transformation as the steel cools. Subsequently, the boltis tempered in the range of 800–1200 degrees F. The actualtempering temperature depends on the bolt chemistry and sizeso that the appropriate hardness range is achieved.Experimental work by Wakiyama (1979) and post-fireexamination of bolts removed from a building by Kirby(1991) and SCI (1991) indicate that, with one exception,  Fig. 3. Representative tensile strength — temperature diagram for structural steel. 28ENGINEERING JOURNAL / FIRST QUARTER / 1998  © 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without the written permission of the publisher.  exposure to fire does not alter high strength bolt properties.When high strength bolts are exposed to fire temperaturesnear the tempering value for 1–2 hours, relaxation of the pre-tensioning due to creep occurs. After the bolt cools toambient temperatures, the srcinal bolt strength is essentiallyregained, Kirby (1986).Because the temperature range between tempering and phase transformation is narrow for high strength bolts, anyexposure to fire above the tempering temperature increasesthe possibility that the bolts have reduced capacity. However,it is highly probable that bolts installed in steel members,which remain undeformed after a fire, have not been exposedto temperatures above the tempering value.Weld metal exposed to elevated fire temperatures can betreated the same as the adjoining base metal when examiningthe metallurgical aspects. The welding process results in aresidual stress magnitude nearly equal to the base metal yieldstrength. This concept is understood for those familiar withwelding and the design specifications account for anylocalized yielding of the steel under service load conditions.The temperature increase resulting from a fire is comparableto post-weld heat treatment commonly used in the pressurevessel industry and addressed by the appropriate ASMEcodes. Chen (1985), Stout (1985) and Zhou (1985) haveaddressed this type of residual stress relief.Wright (1990) examined the post-fire behavior of weldswith tests of welded brackets removed from a bridge having22 years of service. The temperature in the top and bottomflange, based on observed conditions, was predicted at 400degrees F and 1000 degrees F, respectively. This testingrevealed that the bottom flange bracket, which was exposedto a higher fire temperature, had a slightly improved fatiguelife than the top flange bracket. The top flange brackets were  Fig. 4. Representative modulus of elasticity— temperature diagram for structural steel. close to the composite girder neutral axis and thereforesubjected to a very low stress range over the bridge servicelife. Thus, it appears fire-induced stress relieving served tooffset the higher stress range experienced by bottom flange brackets. Representative bridge girder data demonstrating theimproved fatigue life for the bottom flange brackets isreproduced in Figure 5. STEEL BEHAVIOR AT ELEVATEDTEMPERATURES Forces that develop in a steel member, as the temperatureincreases, are primarily dependent on the end restraint, and toa lesser degree on the intermediate bracing and supports.Steel expansion is temperature dependent and as thetemperature of the steel increases, an unrestrained member will elongate according to: ∆  L = α  L ∆ T  where:  L =srcinal length, in. ∆ T  =change in temperature, degrees FA member fully restrained in the axial direction will developaxial stresses as the temperature increases according to:  f  a  =  E  α ∆ T  where:  E = modulus of elasticity, 29,000 ksiTo facilitate future discussion, values of α ,  f  a  and ∆  L  are presented in Table 1 for beam lengths of 25 and 40 ft. Theselengths were chosen because they represent the commonlength range encountered in building construction. Table 1  Fig. 5. Fatigue life of bridge brackets after fire exposure. ENGINEERING JOURNAL / FIRST QUARTER / 1998 29  © 2003 by American Institute of Steel Construction, Inc. All rights reserved. This publication or any part thereof must not be reproduced in any form without the written permission of the publisher.
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